Elsevier Encyclopedia of Geology

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preserved under conditions of incomplete aeration

and high water content. It accumulates in areas

where there is an excess of rainfall and the ground is

poorly drained. Nonetheless, peat deposits tend to be

most common in those regions with a comparatively

cold wet climate. The high water-holding capacity of

peat maintains a surplus of water, which ensures con-

tinued plant growth and consequent peat accumula-

tion. Drying out, groundwater fluctuations, and snow

loading bring about compression in the upper layers

of a peat deposit. As the water table in peat normally

is near the surface, the effective overburden pressure

is negligible.

The void ratio of peat is very large, ranging from 9

up to 25. It usually tends to decrease with depth within

a peat deposit. Such high void ratios give rise to phe-

nomenally high water content, varying from a few

hundreds per cent dry weight to over 3000% in some

coarse fibrous varieties. Put another way, the water

content may range from 75 to 98% by volume of peat.

Peat, therefore, undergoes significant shrinkage on

drying out. The magnitude of pore-water pressure is

particularly significant in determining the stability of

peat. With the exception of those peats with low water

contents (less than 500%) and high mineral contents,

the average bulk density of peat is slightly lower than

that of water. Gas is formed in peat as plant material

decomposes, the volume of gas varying from around 5

to 7.5%. Most of the gas is free and so has a significant

influence on the rate of consolidation, pore pressure

under load and permeability.

Differential and excessive settlement are the princi-

pal problems confronting the engineer working on a

peat soil. Serious shearing stresses are induced, even by

moderate loads. Worse still, should the loads exceed a

given minimum, then settlement may be accompanied

by creep, lateral spread or, in extreme cases, by rota-

tional slip and upheaval of adjacent ground. At any

given time, the total settlement in peat, due to loading,

involves settlement with and without volume change.

Settlement without volume change is the more serious

for it can give rise to the types of failure mentioned

above. What is more, it does not enhance the strength

of peat.

Because of the potential problem of settlement

arising from loading peat, especially in the construc-

tion of embankments carrying roads, some method

of dealing with this problem has to be employed.

Bulk excavation of peat frequently is undertaken if

the deposit is less than 3 m in thickness. When a

deposit exceeds 3 m or peat occurs as layers within

soft sediments, precompression, involving surcharge

loading, commonly is used. With few exceptions, im-

proved drainage has no beneficial effect on the rate of

consolidation.

When peatlands are drained artificially for re-

clamation purposes, the ground level can experience

significant subsidence. The subsidence is not simply

due to the consolidation that occurs as a result of

the loss of the buoyant force of groundwater but

also is attributable to desiccation and shrinkage

associated with drying out in the zone of aeration

and oxidation. For instance, in some parts of the

Fenlands of eastern England, the thickness of peat

has been almost halved as a result of drainage (e.g.,

Holme Post was installed in 1848 and by 1932 the

thickness of the peat had been reduced from 6.7 m to

3.4 m; Figure 9).

Further Reading

Andersland OB and Ladanyi B (1994) An Introduction to

Frozen Ground Engineering. New York: Chapman and

Hall.

Bell FG (1993) Engineering Treatment of Soils. London:

E and FN Spon.

Bell FG (2000) Engineering Properties of Soils and

Rocks, Fourth Edition. Oxford: Blackwell Scientific

Publications.

Blight GE (ed.) (1997) Mechanics of Residual Soils.

Rotterdam: AA Balkema.

Chen FH (1988) Foundations on Expansive Soils.

Amsterdam: Elsevier.

Charman JM (1988) Laterite in Road Pavements. Special

Publication 47, London: Construction Industry Research

and Information Association (CIRIA).

Fookes PG (ed.) (1997) Tropical Residual Soils. Engineering

Group Working Party Revised Report, London:

Geological Society.

Fookes PG and Parry RHG (eds.) (1994) Characteristics of

Arid Soils. Rotterdam: AA Balkema.

Jefferson I, Murray EJ, Faragher E, and Fleming PR (eds.)

(2001) Problematic Soils. London: Thomas Telford Press.

Jefferson IF, Rosenbaum MS, and Smalley IJ (eds.) (2004)

Silt and Siltation, Problems and Engineering Solutions.

Berlin: Springer-Verlag.

ENGINEERING GEOLOGY/Problematic Soils 565

Rock Properties and Their Assessment

F G Bell, British Geological Survey, Keyworth, UK

Introduction

Those properties that can be used to describe rock

materials in terms of engineering classification are

referred to as index properties. Index properties fre-

quently show a good correlation one with another. In

order for the test of an index property to be useful it

is desirable that it should be simple to obtain, rapidly

performed, and inexpensive. The test results must

be reproducible and the index properties must be

relevant to the engineering requirement.

Density and Porosity

The density of a rock is one of its most fundamental

properties. It is influenced principally by its mineral

composition on the one hand and the amount of pore

space on the other; as the proportion of pore space

increases so the density decreases. Four different types

of density are recognised. Firstly, grain density is the

mass of the mineral aggregate per volume of

solid material. The grain density is similar to the spe-

cific gravity (or relative density) of a rock except that

the specific gravity is not expressed in units, it being

the ratio of solid rock to that of an equal volume

of water at a specified temperature. Secondly, dry

density is the mass of the mineral aggregate per volume.

Thirdly, bulk density is the mass of mineral aggregate

and natural water content per volume. Fourthly, sat-

urated density is the mass of mineral aggregate and

saturated water content per volume. The unit weight

can be used instead of density, and is expressed in

terms of stress (kN m

3

). It can be derived from

density simply by dividing by 98.8 (e.g. density of

2200 kg m

3

98.8 ¼ 22.26 kN m

3

unit weight; dens-

ity is usually expressed in Mg m

3

The specific gravity can be determined by grinding

a rock to powder and using a density bottle or alter-

natively by the immersion in water method. Deter-

mination of dry density, bulk density, and saturated

density is dependent upon accurately weighing the

rock specimen concerned and upon the accurate

measurement of its volume. In the case of the dry

density the rock specimen is dried in a ventilated

oven at 105



C until a constant mass is reached and

then allowed to cool in a desiccator. In order to obtain

the saturated density, the rock specimen is first satur-

ated by immersing in water under vacuum with

periodic agitation to remove any trapped air. This

method is not suitable for rocks that slake when

immersed in water. The bulk density simply requires

the mass of the specimen as obtained from the field,

that is, with its natural moisture content. The volume

of specimens can be determined by the caliper method

or the buoyancy method. Regular shaped specimens

are required for the caliper method so that accu-

rate measurements of dimensions are obtained; the

volume is obtained by immersion in water with the

buoyancy method.

The porosity of a rock can be defined as the volume

of the pore space divided by the total volume expres-

sed as a percentage (Figure 1). Grades of dry density

and porosity were suggested by the International

Association of Engineering Geology (IAEG) and are

provided in Table 1. Total or absolute porosity in-

volves the total pore volume, that is, it includes the

occluded pores (occluded pores are isolated and not

interconnected with other pores). However, the ef-

fective or net porosity is a more practical measure-

ment of porosity and it may be regarded as the pore

space from which water or fluid can be removed. The

Figure 1 Different values of porosity attributable to different

degrees of sorting. Numbers beneath diagrams refer to percent-

age porosity.

566 ENGINEERING GEOLOGY/Rock Properties and Their Assessment

latter generally is determined by using the standard

saturation method. Alternatively, an air porosimeter

can be used. In both methods, the pore volume is

obtained by saturating with water or with air, the

total volume being found by the caliper or bouyancy

method. The porosity as determined by these two

tests does not provide an indication of the way in

which the pore space is distributed within a rock, or

whether it consists of many fine pores or a smaller

number of coarse pores. Two tests have been used to

investigate the distribution of pore sizes and the

microporosity of a rock specimen, namely, the suction

plate test and the mercury porosimeter test. Micro-

porosity in the suction plate method is defined as the

volume of water retained (expressed as a percentage

of the total available pore space) when a suction

equivalent to 6.4 m of head of water is applied to

the specimen. In effect, this measures the percentage

of pores with an effective diameter of less than 5 mm.

Such pores are able to retain water against applied

suction and influence the amount of damage that can

be caused by frost or by the crystallization of soluble

salts that a rock used for building stone may undergo.

Mercury is forced to penetrate the pores of the speci-

men under an applied pressure in the mercury poro-

simeter test. Obviously, the finer the pores, the higher

the pressure that must be used to bring about pene-

tration. In this way it is possible to derive the dimen-

sions and pore size distribution from a graph showing

the distribution of pores sizes. A line is drawn on the

curve at the position where 10% of the pore space has

been filled with mercury. The pores below this size

limit can be regarded as the microporosity.

Hardness

Hardness is one of the most investigated properties of

materials, yet it is one of the most complex to under-

stand. It does not lend itself to exact definition in

terms of physical concepts. The numerical value of

hardness is as much a function of the type of test used

as a material property. The concept of hardness is

usually associated with the surface of a material. For

instance, the hardness of a rock can be considered as

its resistance to a penetrating force, whether static or

dynamic, or the resistance to displacement of surface

particles by tangential abrasive force. As such, hard-

ness is controlled by the efficiency of the bond be-

tween minerals or grains, as well as the strength of

these two components.

A number of tests have been used to assess indenta-

tion hardness, of which the two most often used in

rock mechanics testing are the Shore scleroscope and

Schmidt hammer tests. The Shore scleroscope is a non-

destructive hardness measuring device that indicates

the relative values of hardness from the height of

rebound of a small diamond-pointed hammer that is

dropped vertically onto a securely clamped test sur-

face from a height of 250 mm (Figure 2). Because a

rock is not a homogeneous material, several hardness

tests must be made over the surface of the specimen

and the results averaged. Hence, at least 20 hardness

determinations should be taken and each point of test

should be at least 5 mm from any other. The Shore

hardness value can be used to derive an approximate

value of uniaxial compressive strength from Figure 3

for rocks with strengths in excess of 35 MPa.

Figure 2 The Shore scleroscope.

Table 1 Dry density and porosity

Class

Dry density

(Mg m

3

) Description

Porosity

(%) Description

1 Less than 1.8 Very low Over 30 Very high

2 1.8–2.2 Low 30–15 High

3 2.2–2.55 Moderate 15–5 Medium

4 2.55–2.75 High 5–1 Low

5 Over 2.75 Very high Less than

Very low

Reproduced from Bulletin International Association Engineering

Geology, No. 19, 364–371, 1979.

ENGINEERING GEOLOGY/Rock Properties and Their Assessment 567

The Schmidt hammer is a portable non-destructive

device that expends a known amount of stored energy

from a spring and indicates the degree of rebound of a

hammer mass, following impact, within the instrument

(Figure 4). Tests are made by placing the specimen in a

rigid cradle and impacting the hammer at a series of

points along its upper surface. The hammer is held

vertically at right angles to the axis of the specimen.

The specimens should have a flat smooth surface where

tested and the rock beneath this area should be free

from cracks. Test locations should be separated by at

least the diameter of the plunger. At least 20 readings

should be taken from each specimen. The lower 50%

of the test values should be discarded and the average

obtained from the upper 50%. This average is multi-

plied by the correction factor of the Schmidt hammer

to obtain the hardness. However, the Schmidt ham-

mer test is not a satisfactory method for the determin-

ation of the hardness of very soft or very hard rocks.

Like the Shore scleroscope test, Schmidt hardness

values can be used to derive approximate values of

uniaxial compressive strength, as shown in Figure 5.

Abrasion tests measure the resistance of rocks to

wear. The two abrasion tests most frequently used are

the Dorry and Los Angeles tests. As both these tests

are used to assess the resistance to wear of aggregate

for road making, they are not dealt with here.

Figure 3 Correlation chart for Shore hardness, relating unit weight of rock, unconfined compressive strength and hardness value

(After Deere DU and Miller RP (1966) Engineering Classification of Engineering Properties for Intact Rock. Technical Report AFWL-TR-65-116,

Air Force Weapons Laboratory, Kirkland Air Base, New Mexico).

Figure 4 The Schmidt hammer. Reproduced from Engineering

Geology, vol. 4, 1979, Elsevier.

568 ENGINEERING GEOLOGY/Rock Properties and Their Assessment

Deformation of Rocks

The factors that influence the deformation characteris-

tics and the failure of rock can be divided into two

categories: internal and external. The internal category

includes the inherent properties of the rock itself, whilst

the external category is that of its environment at a

particular point in time. As far as the internal factors

are concerned, the mineralogical composition and tex-

ture are obviously important, but fractures within a

rock and the degree of mineral alteration are frequently

more important. The temperature-pressure conditions

under which a rock exists, significantly affect its mech-

anical behaviour, as does its pore water content.

Importantly, the length of time that a rock experiences

a change in stress, and the rate at which this is imposed,

significantly affects its deformation characteristics.

The composition and texture of a rock are gover-

ned by its origin and geological history. Few rocks are

composed of only one mineral species and even when

they are the properties of that species vary slightly

from mineral to mineral. Such variations within min-

erals may be due to cleavage, twinning, inclusions,

cracking, and alteration, as well as to slight differ-

ences in composition. This, in turn, is reflected in the

mechanical behaviour of the resultant rocks. As a

consequence, few rocks can be regarded as homoge-

neous isotropic substances. The size and shape rela-

tionships of the component minerals/grains are also

significant in this respect, generally the smaller the

size, the stronger the rock. One of the most important

features of texture as far as mechanical behaviour,

particularly strength, is concerned, is the degree of

Figure 5 Correlation chart for Schmidt hardness, relating unit weight, unconfined compressive strength and hardness value (After

Deere DU and Miller RP (1966) Engineering Classification of Engineering Properties for Intact Rock. Technical Report AFWL-TR-65-116, Air

Force Weapons Laboratory, Kirkland Air Base, New Mexico).

ENGINEERING GEOLOGY/Rock Properties and Their Assessment 569

interlocking of the component minerals or grains.

Breakage is more likely to take place along grain

boundaries (intergranular fracture) than through

grains (transgranular fracture) and therefore irregular

boundaries make fracture more difficult. The bond

between grains in many sedimentary rocks is pro-

vided by the cement and/or matrix, rather than by

grains interlocking. The amount and, to a lesser

extent, the type of cement/matrix is important, not

only influencing strength and elasticity, but also dens-

ity, porosity, and primary permeability. Rocks are

not uniformly coherent materials, but contain defects

that include microfractures, grain boundaries, min-

eral cleavage, twinning planes, inclusion trains, and

elongated shell fragments. Obviously, such defects

influence the ultimate strength of a rock and may

act as surfaces of weakness that control the direction

in which failure occurs. Crystal grain orientation in

a particular direction facilitates breakage along that

direction.

The presence of moisture in rocks adversely affects

their engineering behaviour. For instance, moisture

content increases the strain velocity and lowers

the strength. More specifically the angle of internal

friction is not affected significantly by changes in

moisture content whereas the cohesion undergoes a

notable reduction. It has therefore been suggested

that the reduction in strength with increasing mois-

ture content is due primarily to a lowering of the

tensile strength, which is a function of the molecular

cohesive strength of the material.

Although all rock types undergo a decrease in

strength with increasing temperature and an increase

in strength as the confining pressure is increased, the

combined effect of these is notably different for dif-

ferent rock types. With increasing temperature there

is a reduction in yield stress and strain hardening

decreases. Heating enhances the ductility of rocks

and their ability to deform permanently without loss

of integrity. The transition from brittle to ductile

deformation in porous rocks is characterized by an

abrupt change from dilational behaviour at low stress

to compaction during inelastic axial strain at high

stress. This type of behaviour differs from that of

rocks with low porosity. With the latter, dilatancy

persists well into the ductile zone. The compaction

that occurs during ductile deformation in porous

rocks at high confining stress is due to collapse of

the pore space and the rearrangement of grains to

give more compact packing.

Four stages of deformation have been recognized,

namely: elastic, elastico-viscous, plastic, and rupture.

The stages are dependent upon the elasticity, viscosity,

and rigidity of the rock, as well as on stress history,

temperature, time, pore water, and anisotropy. An

elastic deformation is defined as one that disappears

when the stress responsible for it ceases. Ideal elasti-

city exists if the deformation on loading and its disap-

pearance on unloading are both instantaneous. This is

never the case with rocks since there is always some

retardation, known as hysteresis, in the unloading

process. With purely elastic deformation the strain is

a linear function of stress, that is, the material obeys

Hooke’s law. Therefore, the relationship between

stress and strain is constant, and is referred to as

Young’s modulus, E. Rock only approximates to an

ideal Hookean solid. In fact, Young’s modulus is not a

simple constant but is related to the level of applied

stress. Just how closely rock approximates to an ideal

material depends on its homogeneity, isotropy, and

continuity. Homogeneity refers to the physical con-

tinuity of a material, that is, the constituent particles

are evenly distributed throughout its volume so that

the elastic properties are the same at all points. Isot-

ropy represents a measure of the directional properties

of a rock. Hence, a rock is only isotropic if it is mono-

mineralic and the crystals/grains have a random orien-

tation. Since most rocks are composed of two or more

essential minerals, which may possess preferred orien-

tation, they are generally anisotropic. Continuity

refers to the pore space and fractures within a rock.

The degree of continuity affects the cohesion and so

the transmission of stress throughout a rock.

The change in deformability at the elastic limit

from elastic to plastic deformation is referred to as

the yield point or yield strength. If the stress acting on

a rock exceeds its elastic limit, then it becomes per-

manently strained, the latter being brought about by

plastic flow. Within the zone of plastic flow there is a

region where elastic stress is still important, referred

to as the field of elastico-viscous flow. Plasticity

may be regarded as time-independent, non-elastic,

non-recoverable, stress-dependent deformation under

uniform sustained load. Solids are classified as brittle

or ductile according to the amount of plastic deform-

ation they exhibit. In brittle materials the amount of

plastic deformation is zero or very little, whereas it is

large in ductile substances. Rupture occurs when the

stress exceeds the strength of the material involved.

It represents the maximum stress a rock is able to

withstand prior to loss of cohesion by fracturing.

The initiation of rupture is marked by an increase in

strain velocity.

Most strong rocks exhibit little time-dependent

strain or creep. However, creep in evaporitic rocks,

notably salt, may greatly exceed the instantaneous

elastic deformation. The time-strain pattern exhibi-

ted by such rocks, when subjected to a constant uni-

axial stress, can be represented diagramatically as

shown in Figure 6. The instantaneous elastic strain,

570 ENGINEERING GEOLOGY/Rock Properties and Their Assessment

which takes place when a load is applied, is repre-

sented by OA. There follows a period of primary or

transient creep (AB) in which the rate of deforma-

tion decreases with time. If the stress is removed,

the specimen recovers. At first this is instantaneous

(BC), but this is followed by a time elastic recovery,

illustrated by curve CD. On the other hand, if

the loading continues, the specimen begins to exhibit

secondary or pseudo-viscous creep. This type of creep

represents a phase of deformation in which the rate of

strain is constant. The deformation is permanent and

is proportional to the length of time over which the

stress is applied. If the loading is continued further,

then the specimen suffers tertiary creep during which

the strain rate accelerates with time and ultimately

leads to failure.

Elastic Properties

Young’s modulus, E, is the most important of the

elastic constants and can be derived from the slope

of the strain–stress curve obtained when a rock speci-

men is subjected to unconfined compression (i.e.,

static loading), it being the ratio of stress to strain.

The strains are measured by attaching strain gauges

to the test specimens, or by displacement transducers,

and recording their outputs. Strain measurements on

specimens less than 50 mm in diameter, however, are

high and not representative of the material behaviour.

Most crystalline rocks have S-shaped stress–strain

curves (Figure 7). At low stresses the curve is non-

linear and concave upwards, that is, Young’s modulus

increases as the stress increases. The initial tangent

modulus is given by the slope of the stress-strain curve

at the origin. Gradually a level of stress is reached

where the slope of the curve becomes approximately

linear. In this region Young’s modulus is defined as

the tangent modulus or secant modulus. At this stress

level the secant modulus has a lower value than

the tangent modulus because it includes the initial

‘plastic’ history of the curve. A classification of

deformability has been suggested by the IAEG and is

given in Table 2.

In addition to their non-elastic behaviour, most

rocks exhibit hysteresis. Under uniaxial stress the

slope of the stress–strain curve during unloading ini-

tially is greater than during loading for all stress

values (Figure 7). As stress is decreased to zero a

residual strain, OR, is often exhibited. On reloading

the curve RS is produced that, in turn, is somewhat

steeper than OP. Further cycles of unloading and

reloading to the same maximum stress give rise to

hysteresis loops, which are shifted slightly to the

right. The non-linear elastic behaviour and elastic

hysteresis of brittle rocks under uniaxial compression

is due to the presence of flaws or minute cracks in the

rock. At low stresses these cracks are open but they

close as the stress is increased and the rock becomes

elastically stiffer, that is, E increases with stress. Once

the cracks are closed the stress–strain curve becomes

linear.

When a specimen undergoes compression it is

shortened and this generally is accompanied by an

increase in its cross-sectional area. The ratio of lateral

Figure 7 Representative stress–strain curve for rock in uncon-

fined compression, showing hysteresis.

Table 2 Classification of deformability

Class Deformability (MPa  10

3

) Description

1 Less than 5 Very high

2 5–15 High

3 15–30 Moderate

4 30–60 Low

5 Over 60 Very low

Reproduced from Bulletin International Association Engineering

Geology, No. 19, 364–371, 1979.

Figure 6 Theoretical strain curve at constant stress (creep

curve).

ENGINEERING GEOLOGY/Rock Properties and Their Assessment 571

unit deformation to linear unit deformation, with-

in the elastic range, is known as Poisson’s ratio, n.

This similarly can be obtained by monitoring strains

during an unconfined compression test. The ideal

geometrical value of Poisson’s ratio is 0.333.

Another elastic constant is compressibility, K,

which is the ratio of change in volume of an elastic

solid to change in hydrostatic pressure. A further

measure of elasticity is rigidity, G, which refers to

the resistance of a body to shear. These four elastic

constants (E, n, K, G) are not independent of each

other and, if any two are known, it is possible to

derive the other two from the following expressions:

G ¼ E=2ð1 þ nÞ½1

and

K ¼ E=3ð1  2nÞ½2

Of the four constants, Young’s modulus and Poisson’s

ratio are more readily determined experimentally.

Methods used to determine the dynamic values (as

opposed to static values, see above) of Young’s modu-

lus and Poisson’s ratio generally depend upon deter-

mining the velocities of propagation of elastic waves

through a specimen of rock. These can be measured

by using the high frequency ultrasonic pulse method,

the low-frequency ultrasonic pulse technique, or the

resonant method. For example, the high-frequency

ultrasonic pulse method is used to determine the

velocities of compressional, v

, and shear, v

, waves

in rock specimens of effectively infinite extent com-

pared to the wavelength of the pulse used. The condi-

tion of infinite extent is satisfied if the average grain

size is less than the wavelength of the pulse that, in

turn, is less than the minimum dimensions of the

specimen. These two velocities can be substituted in

the following expressions to derive the dynamic

values of Young’s modulus and Poisson’s ratio:

E ¼ rn

ð1 þ nÞð1  2nÞ

ð1  nÞ

½3

E ¼ 2n

rð1  nÞ½4

E ¼ n

½3ðn

 4

½ðn

 1

½5

n ¼

0:5ðn

 1

ðn

 1

½6

where r is density.

Strength

Uniaxial Compression

The uniaxial strength, also known as the unconfined

compressive strength, of a rock may be regarded as

the highest stress that a rock specimen can carry when

a unidirectional stress is applied, normally in an axial

direction to the ends of a cylindrical specimen. It

represents the maximum load supported by a speci-

men during the test divided by the cross-sectional area

of the specimen. Grades of unconfined compressive

strength are shown in Table 3. Although its applica-

tion is limited, the uniaxial compressive strength

allows comparisons to be made between rocks and

affords some indication of rock behaviour under more

complex stress systems.

The behaviour of rock in uniaxial compression is

influenced to some extent by the test conditions. The

most important of these is the length-diameter or

slenderness ratio of the specimen, the most satisfac-

tory slenderness ratio being 2.5 since it provides a

reasonably good distribution of stress throughout

Table 3 Grades of unconfined compressive strength

Geological Society (1977) IAEG (1979)



ISRM (1981)



Term Strength (MPa) Term Strength (MPa) Term Strength (MPa)

Very weak Less than 1.25 Weak Under 15 Very low Under 6

Weak 1.25–5.00 Moderately strong 15–50 Low 6–10

Moderately weak 5.00–12.50 Strong 50–120 Moderate 20–60

Moderately strong 12.50–50 Very strong 120–230 High 60–200

Strong 50–100 Extremely strong Over 230 Very high Over 200

Very strong 100–200

Extremely strong Over 200



International Association of Engineering Geology.



International Society for Rock Mechanics.

572 ENGINEERING GEOLOGY/Rock Properties and Their Assessment

the specimen. Secondly, the rate at which loading

occurs also influences the compressive strength, a

loading rate of between 0.5 and 1.0 MPa normally is

recommended. Thirdly, the ends of a specimen should

be lapped so that they are exactly perpendicular to the

long axis.

The onset of failure in a rock specimen subjected to

compressive loading is marked initially by the forma-

tion of a large number of isolated fractures, which

characterizes the relief of stress concentration pro-

duced by the mechanical inhomogeneities in the

rock. Most cracks are orientated parallel to the ap-

plied stress (Figure 8). This is quickly followed by the

development of two groups of macroscopic shear

failures, at the boundary and in the interior of the

specimen, which suggests that most of the major

sources of induced lateral tensile stresses have been

eliminated. The interior macroscopic shear failures

are extended and become interconnected to form a

conjugate set of open shear fractures. Two central

cones are formed, which either abrade during the

large shear displacement or produce major fractures

in the remaining rock material. If one shear failure

surface becomes dominant, then cones are not de-

veloped and the sample ultimately fails in two parts

along a diagonal plane of shear.

Triaxial Compression Strength

A triaxial test is necessary if the complete nature of

the failure of a rock is required. In this test a constant

hydraulic pressure (the confining pressure) is applied

to the cylindrical surface of the rock specimen, whilst

applying an axial load to the ends of the sample. The

axial load is increased up to the point where the

specimen fails. Testing of the rock specimen is carried

out within a special high pressure cell (Figure 9).

A series of tests, each at higher confining pressure,

are carried out on specimens from the same rock.

These enable Mohr circles and their envelope to be

drawn, from which the strength parameters, that is,

the angle of friction (f) and value of the cohesion (c)

are obtained (Figure 10). The shear strength (t) is then

derived from the Mohr–Coulomb criterion:

t ¼ c þ s tanf ½7

where s is the normal stress.

Direct Shear

A number of other tests can be used to deduce the

shear strength of intact rock and of rock fracture

surfaces, in addition to the triaxial test. These include

the shear box test, the direct single and double shear

tests, the punch shear test, and the torsion test. The

most commonly used method is the shear box test,

which is particularly useful for assessing the shear

strength of weaker rocks and also for obtaining the

shear strength along discontinuity surfaces. In this

test a constant normal force is applied to the speci-

men, which then is sheared. A number of tests on the

same material are undertaken, each at a higher normal

stress, so that a shear strength versus normal stress

graph can be drawn from which the value of cohesion

and angle of friction are derived (Figure 11A). The

other shear tests are used infrequently but are illus-

trated in Figure 11B.

Tensile Strength

Rocks have a much lower tensile strength than com-

pressive strength. Brittle failure theory predicts a ratio

of compressive strength to tensile strength of about

8:1 but in practice it is generally between 15:1 and

25:1. The direct tensile strength of a rock can be

obtained by attaching metal end caps with epoxy

resin to the specimen, which are then pulled into

tension by wires. In direct tensile tests the slenderness

ratio of cylindrical specimens should be 2.5 to 3.0 and

the diameter preferably should not be less than

54 mm. The ratio of the diameter of the specimen to

the largest crystal/grain in the rock should be at least

10:1. Unfortunately, the determination of the direct

tensile strength often proves difficult since a satisfac-

tory method has not yet been devised to grip the

specimen without introducing bending stresses. Ac-

cordingly, most tensile tests have been carried out by

indirect methods.

The Brazilian test is an indirect method of asses-

sing the tensile strength of rocks, based on the obser-

vation that most rocks in biaxial stress fields fail in

tension when one principal stress is compressive. In

this test a cylindrical specimen of rock is loaded in a

diametrical plane along its axis. The sample usually

fails by splitting along the line of diametrical loading

and the indirect tensile strength (T

) can be obtained

from:

Figure 8 Stages in the development of fracturing with

increasing unconfined compressive loading.

ENGINEERING GEOLOGY/Rock Properties and Their Assessment 573

¼ 2P=pLD ½8

where P is the load at failure, and L and D are the

length and diameter of the specimen respectively.

Disc-shaped specimens are used in the Brazilian

disc test. In this case curved jaw loading platens are

used to improve loading conditions. Uncertainties

associated with the premature development of failure

can be removed by drilling a hole in the centre of the

disc-shaped specimen (sometimes this has been re-

ferred to as a ring test). A disc-shaped specimen

should be wrapped around its periphery with a layer

of masking tape and the specimen should not be less

Figure 9 Cutaway view of a triaxial cell for testing rock.

Figure 10 Mohr envelope and circles.

574 ENGINEERING GEOLOGY/Rock Properties and Their Assessment

Elsevier Encyclopedia of Geology - vol I A-E

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